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analysis of steel concrete composite beam to column jonits bolted solutions


ANALYSIS OF STEEL-CONCRETE COMPOSITE BEAM-TO-COLUMN JOINTS: BOLTED SOLUTIONS
Oreste S. Bursi Department of Mechanical and Structural Engineering, University of Trento Trento, Italy oreste.bursi@ing.unitn.it Fabio Ferrario Department of Mechanical and Structural Engineering, University of Trento Trento, Italy fabio.ferrario@ing.unitn.it Raffaele Pucinotti Department of Mechanics and Materials, Mediterranean University of Reggio Calabria Reggio Calabria, Italy raffaele.pucinotti@unirc.it Riccardo Zandonini Department of Mechanical and Structural Engineering, University of Trento Trento, Italy riccardo.zandonini@ing.unitn.it

ABSTRACT In this paper, a multi-objective advanced design methodology is proposed for steel-concrete composite moment-resisting frames. The research activity mainly focused on the design of the beam-to-column joints under seismic-induced fire loading together with the definition of adequate structural details for composite columns. Thermal analyses of cross sections were performed in order to obtain internal temperature distribution; structural analyses were then performed on the whole frame to assess the global behavior under the combined action of static and fire loadings. Besides, results of numerical analyses were used in order to derive information about the mechanical and numerical behavior of joints. In this paper the experimental program carried out on four beam-to-column joint specimens are described and experimental results are presented and discussed together with the outcomes of numerical simulations. Experimental tests demonstrated the adequacy of the seismic design. Numerical simulations showed a satisfactory performance of joints under seismic-induced fire loading. INTRODUCTION Steel-concrete composite structures are becoming increasingly popular around the world due to the favorable stiffness, strength and ductility performance of composite systems under seismic loading, and also due to the speed and ease of erection. Moreover, such structures exhibit better fire resistance characteristics compared to bare steel structures. Considering the high

probability of a fire after a seismic event, the use of steel–concrete composite structures in seismic areas, potentially represents a fairly effective design solution. In fact, the adoption of such a solution is in general more efficient from both a structural and a constructional viewpoint when compared to bare steel structures. The benefit increases, if the high probability that an earthquake and a fire can occur in sequence. In current Eurocodes for structural design, seismic and fire safety are accounted for separately, and no sequence of seismic and fire loading is taken into account. In reality, the risk of loss of life increases if a fire occurs in a building after an earthquake. In the Kobe Earthquake (1995) many people died due to the collapse of buildings exposed to a fire that followed an earthquake; in fact, large sections of the city burned, greatly contributing to the loss of lives. It is obvious therefore that seismic-induced fire is a design scenario that should be properly addressed in any performance-based seismic design. In this situation the traditional single-objective design is not adequate, and a multiobjective advanced design has to be adopted. This enables a proper account of: (i) seismic safety with regard to accidental actions; (ii) fire safety with regard to accidental actions (iii) fire safety on a structure characterized by stiffness deterioration and strength degradation due to seismic actions. As a result, the fire design applied to a structure with reduced capacity due to seismic damage will permit simultaneous fulfillment of the requirements associated with structural, seismic and fire safety ,and with structural fire safety and structural seismic safety separately taken as accidental actions. In order to characterize the seismic behaviour of composite joints, a research project has been carried out by the University of Trento with the objective of developing a design procedure for composite joints under the ‘combined’ action of earthquake and post-earthquake fire. The research activity was mainly concerned with the design of bolted beam-to-column joints with CFT columns with circular hollow steel sections. This solution derives from a parent welded design solution developed in an European project (Bursi et al. 2006) and is aimed at ensuring easiness of assembly and avoiding the problems related with on site welding. The results of the experimental programme devoted to the evaluation of the cyclic behavior of beam-to-column bolted joints are reported in the following. In order to better understand the activation of all the transfer mechanisms proposed in the Eurocode 8 (UNI EN 1998-1, 2005), a numerical finite element (FE) model of the slab has been developed, together with FE model of the joint under fire. SEISMIC AND FIRE DESIGN OF REFERENCE FRAMES A reference composite steel-concrete office-building was considered, made up of three five storeys moment resisting frames, placed at the distance of 7.5 m each in the longitudinal direction and braced in the transverse direction. The storey height was equal to 3.5 m. Two moment resisting frames, having the same structural typology but different slab systems, were analyzed: i) a composite steel-concrete slab with structural profiled ribbed steel sheeting; ii) a concrete slab composed of electro-welded lattice girders. Since the two slab systems had different load bearing capacities, a different distance between the secondary beams was adopted for the two solutions. The distance slightly affected the frame geometry. The first solution (steel sheeting slab), is depicted in Figure 1a while the second one, with the slab on prefabricated lattice elements, is shown in Figure 1b. The elevation is shown in Figure 1c (Bursi et. al 2008). Two different types of composite beams were hence designed according to Eurocode 4 (UNI EN 1994-1-1, 2005); the steel section was maintained in both cases an IPE400 of steel grade S355. The slab, 150 mm deep, was designed in accordance with the relevant specifications of Section 9 of Eurocode; moreover design procedure indicated by the producer of the prefabricated lattice elements was followed in the second case. All connections between the steel beam and the slab were made by Nelson 19 mm stud connectors made of steel with an ultimate tensile strength fu=450 MPa.

a) b) c) Figure 1 - The configuration of the building and the longitudinal five-storey MR Frame; slab with a) profiled ribbed steel sheeting; b) slab with prefabricated lattice girder; c) elevation The frames were designed to have a dissipative structural behavior according to Eurocode 8 (UNI EN 1998-1, 2005). In detail two following structural types were considered. 1. Moment Resisting Frames: in which dissipative zones were mainly located to the end of the beams, in the vicinity of beam-to-column connections, where plastic hinges could develop (UNI EN 1998-1, 2005, 7.1.2 b); 2. Concentrically braced frames: in which the horizontal forces were mainly resisted by members subjected to axial forces with dissipative zones located on the tension diagonals only (UNI EN 1998-1, 2005, 7.1.2 c). The effective width of the slab in composite beams was determined according to Eurocode 4 (UNI EN 1994-1-1, 2005, 5.4.1.2) for static and fire analyses, and to Eurocode 8 (UNI EN 19981, 2005, 7.6.3) for seismic analyses. Beam-to-column connections resulted to be rigid according to Eurocode criterion (UNI EN 1994-1-1, 5.1.2). The design of structural elements were performed considering static and seismic design situations. Moreover, numerical simulations on two-dimensional (2D) frames, were first performed by means of the SAFIR program (Franssen J.-M. 2000), in order to study different fire scenarios acting in the reference buildings and to evaluate the performance of different elements, i.e. composite beams, composite columns, and beam-to-column joints, under fire load for different times of exposure to fire. Five fire scenarios were considered. In the first one (FS1), fire acted only into a span in the first floor as illustrated in Figure 3a.

a) FS1

b) FS2

c) FS3

d) FS4

e) FS5

Figure 3 - Fire scenarios considered in thermal analysis In the second one (FS2), fire acted on the whole first floor; this means that the first level columns and the first level beams were heated. In the third one (FS3), fire acted only into a span in the last floor. In the fourth one (FS4), fire acted on the whole fifth floor. Finally, in the fifth one (FS5) fire acted on the whole frame. The fire load followed the ISO 834 fire law. Figure 4 shows the evolution of the bending moment and of the axial force as a function of time at various locations in the FE model for the case FS1. In detail the bending moment in the column

C shows an inversion in the sign when axial forces in the beams changes sign too. In fact, the increase of the temperature cause first an increment of axial load in the beam (compression) until to about 18 min owing the presence of the column restraint. Afterwards, the reduction of stiffness of columns subject to fire prevails and the axial force in the beam turns to tension, being similar to a “catenary” structure (with large displacement and deflection). The elongation of the beam due to increase in temperature and the different “restraint” effect offered by the column also causes a reversal of the sign of bending moment at mid-span of the beam which from sagging becomes hogging and then sagging again. The results of the analysis shows that the frame collapses because of the formation of a beam mechanism in the longest span involving the formation of three plastic hinges located at mid-span and at both beam ends near supports. Figure 5 shows column temperature distributions in the fire case FS1; it is possible to notice that the steel tubular section offers a practically negligible protection to internal concrete; therefore temperatures of steel elements directly exposed to fire are equal to that of the external atmosphere.
Column C - first floor
1000.00 500.00

M (kNm)

0.00 0 -500.00 5 10 15 20 25 30

151 155 160 161

165

170 110

-1000.00

A

B

C
101

-1500.00

elem. 101 elem. 110

Time (min)

Beams A & B - first Floor
600 400 400.00

Beam B - first floor

Axial Load (kN)

M (kNm)

200 0 5 10 15 20 25 30

0.00 0 -400.00 elem. 161 elem. 165 elem. 170 5 10 15 20 25 30

-200 0 -400 -600
compressione

elem. 165 elem. 155 -800.00

-800

Time (min)

Time (min)

Figure 4 – Fire Case FS1: Bending Moment and Axial Load in beams and columns of the first storey

10’

30’

40’

Figure 5 - Temperature distributions inside the column at different time. Fire Case FS1 The thermal distribution, for the composite slab with steel sheeting, evaluated after 10 min, 30 min and 40 min of ISO fire, by means of the thermal FE model created by SAFIR are presented in Figure 6, while Figure 7 shows the thermal distribution for the case of composite slab with prefabricated slab. As expected, the temperature increment was higher in the composite slab with steel sheeting than in the slab with prefabricated R.C. elements. As a result, the moment resistance of the former slab was lower than in the latter owing to the higher reduction in materials strength.

10’

30’

40’

Figure 6 – Fire case FS1. Evolution of the temperature distribution inside beam B with a steel sheeting composite slab.

10’

30’

40’

Figure 7 – Fire case FS1. Evolution of temperature distribution inside beam B with prefabricated elements BEAM-TO-COLUMN JOINTS DESIGN The joint design aimed at ensuring to the joint between the column and the beam an overstrength with respect to the beam. The proposed bolted solution derives from a parent welded design (Bursi et. al 2008), and was conceived to guarantee easiness of assembly and to avoid problems related to welding on site. The joint is comprised of two horizontal diaphragm plates and a vertical through-column plate attached on the tube by groove welds as indicated in Figure 8.
2x HORIZONTAL PLATES VERTICAL PLATE
23 °

23°

23°

23°
23 °

16

18 2,3

168

= 320 x 1094 x 14
16
18

36 124 36

75 75 75

1094 572

75 75 75

36

14

18

,5 28 R2

= 660 x 1304 x 16
320 1304 502 300 502

Column

= 320 x 1094 x 14
230

36

124

Section A-A
46
1304 300

A

A

A

a=8 L=320

A

66 66 66 66 66 36 660 200 46 108

a=8 L=320
,5 28 R2

230

17

14

a=8 L=1475

660

18

166

= 660 x 1304 x 16

4,3

a=8 L=320

a=8 L=320

Figure 8 - Steel-concrete composite beam-to-column joint: a) horizontal and vertical plates; b) Nelson studs in columns The flanges and the web of each beam were connected to the horizontal plates and the vertical plate by two and three rows of bolts M27 10.9, respectively. A part the slab type, joints differed owing to the presence of strengthening plates with holes required by Eurocode 8 to avoid brittle fracture in net sections of bolted joints. Thus specimens JB-P1 and JB-S1, see table 1, where endowed with extra plates; whereas JB-P2 and JB-S2 had no plates being beam-to-column joints not dissipative. Beam-to-column joint design was developed using the component method. The joint was simulated by a series of different components in agreement with Eurocode 3 par 1-8 and Eurocode 8 (UNI EN 1993-1-8, 2005, UNI EN 1998-1, 2005), achieving the necessary overstrength to the joint in respect to the connected composite beams.

17

Stiffness and strength of complex components, like top and bottom plates or concrete slab in compression, were defined by means of refined Finite Element (FE) models of the joint including friction between the slab and the column was developed. Depending on the level of friction, the distribution of the compression forces in the slab under sagging bending moment was found to be different: localized in front of the column for a friction coefficient equal to 0,35 as indicated in Figure 9a; while it spreads over a more extended portion of the slab for increasing values of the friction coefficient. In detail, the diffusion angle becomes greater than 80 degrees for a friction coefficient of about 1. The results show that, in order to activate the transfer mechanisms proposed in the Eurocode 8 (UNI EN 1998-1, 2005), i.e. Mechanism 1 front mechanism and Mechanism 2 strut and tie mechanism, it is necessary to increase the level of friction between the concrete slab and the composite column. As a result, Nelson 19 mm stud connectors welded around the column were used in the tested specimens as indicated in Figure 8. As a results beam-to-column joints were rigid full-strength joints satisfying the relation: (1) M j , Rd ≥ s ? γ ov ? M b , pl , Rd in which Mj,Rd is the resisting moments of full-strength beam-to-column joints and Mb,pl,Rd is the resisting moment of the composite beam (UNI EN 1998-1, 2005). Moreover, the ductile behaviour of joints was guaranteed by the following relationships: (2) Rd ,bolt ≥ s ? γ oν ? R pl ,Rd ,beam

Rd ,bolt ≥ R pl ,Rd , plates
Fv ,Rd ≥ Fb ,Rd
1
40 °

(3) (4)

2

2

80° 80°

° 4 40

a)

b)

Figure 9 - Distribution of compression stresses in the slab for: (a) friction coefficient equal to 0,35; b) friction coefficient equal to 1 where s = min f t f y ;1.25 , γ oν = 1.1 , Rd ,bolt is the design strength of bolts, R pl ,Rd ,beam is the plastic design strength of the beam, R pl ,Rd , plates is the plastic design strength of the plates, while

{

}

Fv ,Rd and Fb ,Rd are the design shear strength and design bearing resistance of bolts,
respectively. The following conditions were also fulfilled for joints JB-P1 and JB-S1 to avoid brittle fracture, i.e.:

0.90 Anet f u

γM2



Af y

γM0

and

0.90 Anet f u

γM2



Af f y

γM0

(5)

where A f is the area of the tension flange. Nevertheless, being joints not dissipative, JB-P2 an JB-S2 did not. A 3D finite model of the interior joint was implemented in the Abaqus 6.4.1 code (Hibbitt et al. 2000). Then, fire design of beam-to-column joints subjected to fire load were conducted. In particular, the component approach, exclusively applied to the moment-rotationtemperature behaviour, in the absence of axial thrust owing to the thermal expansion restraint of the beam, was adopted, and a simplified model was derived, which can predict the moment-

rotation-temperature characteristic of the joint. As a results, at high temperatures, the joint were designed to transfer shear forces owing to vertical loads from a beam to the other. Accordingly, vertical through column plate, top horizontal plate together with Nelson stud connectors welded around the column were arranged. In addition, two longitudinal steel rebars were added to reduce the damage produced by the seismic actions before fire. EXPERIMENTAL TESTS The experimental programme consisted of 4 tests under cyclic loading of full-scale substructures representing interior full-strength bolted beam-to-column joint (Figure 10). Experimental tests were carried out at the Laboratory for Materials and Structures of the University of Trento. Joint specimens were subjected to cyclic loadings up to collapse, according to the ECCS stepwise increasing amplitude loading protocol, modified with the SAC procedure (ECCS 1986, SAC 1997) by using ey=0.005h=17.5 mm where h represents the storey height.
5700 2850 280 30
( 310 + 241.3 )

2850 5140 500 280 2320
280 30
( 310 + 241.3 )

5700 2850 2320 2290 5140 500 2850 280 2320

2320 2290

550

1750 1200

A

1750 1200

550

A

A

1344
CFT ? 457 x 12

2590 2670

A

1344
CFT ? 457 x 12

1750 1470

1470

a)

1230

b)

40

Figure 10: Bolted beam-to-column specimens a) slab with electro-welded lattice girders, b) composite slab with profiled steel sheeting The slab reinforcement, in the composite steel-concrete beams with steel sheeting, consisted of 3+3φ12 longitudinal steel rebars in order to carry the sagging moment, and of 4+4φ12@100 mm and 7+7φ16@250 mm transverse steel rebars, in order to enable development of the seismic slab-to-column transfer mechanism as well as the resistance to the shear force. A mesh φ6@200x200 mm is also present as highlighted in Figure 11a. The concrete class was C30/37 while the steel grade S450 was adopted for reinforcing steel bars. In the case of the concrete slab prefabricated R.C. elements, the slab reinforcement was made up of 3+3φ12 longitudinal steel rebars and by 5+5φ12@100 mm plus 8+8φ16@200 mm transverse steel rebars. The same mesh was adopted as for the composite slab (Figure 11b).
pos.A 4 ? 12 / 100 mesh ? 6 / 200x200
160 300150 1500

80

pos.A 5 ? 12 / 100

160

400

150

1400

pos.B 3+3 ? 12

mesh HD 620

3?12

pos.B 3+3 ? 12

a)

b)

Figure 11 – Rehears layout in: a) steel sheeting slab; b) prefabricate lattice girder slab

1?8 pos.E HD 10/14/8 h=9,5cm 3?12 pos.B mesh HD 620 1?10 pos.H 1?10 1?10 1?10 pos.H pos.D pos.D

pos.C 7 ? 16 / 250

mesh ? 6 / 20x200

pos.C 8 ? 16 / 200

2000

2000

560 720 880

3?12

560 720 880

40

1230

2590 2670

IPE400

IPE400

IPE400

IPE400

The columns were concrete-filled columns with a circular hollow steel section with a diameter of 457 mm and a thickness of 12 mm. Steel grade is S355. The column reinforcement consisted of 8φ16 longitudinal steel rebars and of stirrups φ8@150 mm (Figure 12). The concrete class was C30/37, while the steel grade S450 was adopted for the reinforcing steel bars. Actual values better than nominal ones can be found in (Bursi et al. 2008). A scheme of the test set-up is shown in Figure 13.
pos.E ? 8 / 15 cm

325 135

40

50 120 150 150 150

pos.D 8 ? 16

2490

90

95

2670

Figure 12 - Column and column reinforcement
SLAB
I INCLINOMETERS L LVDTs
4

SG STRAING GAUGE O OMEGA STRAING GAUGE LVDTs L SG L L L O SGL SG OO L L SG SG L L SG L O

SG

L L L I I L L

L I I

INFERIOR PLATE
SG

BEAM

SG SG

SG SG

BEAM
SG

SG

SUPERIOR PLATE

SG

Figure 13 - Lateral view of the Test set-up and main instrumentation on specimens No vertical actuator was imposed on the column to be consistent with tests in other labs. In the tests, the following instrumentation were employed as illustrated in Figure 13: a) 5 inclinometers measured the inclinations: of the column at the top and, in the zone adjacent to the joint, and of the beams near the connection; b) 4 LVDTs detected the interface slip between the steel beam and the concrete slab and between the bottom horizontal joint plate and the flange of beam; c) 2 LVDTs were employed in order to measure the bottom horizontal joint plate deformations; d) 10 LVDTs were utilized in order to measure concrete slab deformations in the zone around the column; e) 4 ? strain gauges detected the deformations of the concrete slab; e) 8 strain gauges monitored axial deformations of the reinforcing bars to scrutinise the effective breadth of the reinforcing bars at each loading stage; f) 4 strain gauges monitored deformations of top and bottom horizontal plates; g) 4 strain gauges recorded flange strains in order to estimate internal forces in steel beams; h) 2 load cell were set on the top of pendula and were utilized in order to measure the horizontal and vertical components of the forces; i) 1 digital transducer (Heidenhein DT500) in order to measure the top column displacement. RESULTS AND SIMULATIONS The observed response clearly indicated that all specimens exhibit a good performance in terms of resistance, stiffness, energy dissipation and local ductility. Both the overall forcedisplacement relationship and the moment-rotation relationships relevant to plastic hinges formed in composite beams exhibited a hysteretic behaviour with large energy dissipation without evident loss of resistance and stiffness as indicated in Figures 14-16. Hysteretic loops of moment-rotation relationship are unsymmetrical owing the different flexural resistance of the composite beam under hogging and sagging moments as can be observed in Figure 16. It was evident that owing to local buckling effects, severe strength degradations appeared to exceed

20 per cent of maximum strength values with rotations of about 40 mrad. The collapse of all specimens was due to fracture of the bottom beam flange near the horizontal plate as illustrated in Figure 17.
Specimen JB-P1
1000 750

Specimen JB-P2
1000 750 500

Force [kN]

Force [kN]

500 250 0 -14 -10 -6 -2 -250 -500 -750 -1000 2 6 10 14

250 -14 -10 -6 0 -2 -250 -500 -750 -1000 2 6 10 14

Displacement [e/ey]

Displacement [e/ey]

Figure 14 - Specimens JB-P1 and JB-P2 – Force-Displacement relationship
Specimen JB-S1
1000 750 500 250 0 -14 -10 -6 -2 -250 -500 -750 -1000 2 6 10 14

Specimen JB-S2
1000 750 500

Force [kN]

Force[kN]

250 -10 -6 0 -2 -250 -500 -750 -1000 2 6 10 14

-14

Displacement [e/ey]

Displacement [e/ey]

Figure 15 - Specimens JB-S1 and JB-S2 - Force-Displacement relationship
Specimen JB-S1
1000 750 500

M [kNm]

250 -60 -40 0 -20 -250 0 -500 -750 -1000 M (plastic hinge at dx) M (plastic hinge at sx) 20 40 60 80

-80

φ [mrad]

Figure 16 - Moment-rotation relationship of the plastic hinges for the JB-S1

Figure 17 – Specimen JB-S1, plastic hinge in the composite beam

Table 1 reports some key experimental data for each test: maximum applied displacement d, maximum value F of the force, maximum values of both sagging and hogging moment M, total number Ntot of cycles performed during the tests. Joints behaviour is essentially symmetric in terms of force, while, for all specimens, sagging moments are about 40% larger of hogging moments. Figure 18 shows a comparison between Force-Interstorey drift curves for the specimens with and without plate strengthening of the horizontal diaphragms in both prefabricated slab and steel sheeting slab. These reinforcing plates were introduced in order to satisfy the relationships in (5). In this case, in which dissipative zones were located to the end of the beams, beam-to-column joints were rigid full-strength and the introduced reinforcing plates did not influence the specimen behaviour. Moreover, beam-to-column joints subjected to fire load were analyzed. In particular, the component approach, exclusively applied to the moment-rotation-temperature behaviour, in the absence of axial thrust owing to the thermal expansion restraint of the beam, was adopted, and a simplified model was derived, which can predict the moment-rotation-temperature characteristic of the joint. The method is capable of predicting the variation of failure modes owing to change in the joint geometrical and material properties, as well as loading conditions.

JB-P1 vs. JB-P2
800 BJ-P1 BJ-P2 400 0 -8 -6 -4 -2 -400 -800 0 2 4 6 8
BJ-S1 BJ-S2

JB-S1 vs. JB-S2
800 400 0 -8 -6 -4 -2 -400 -800 0 2 4 6 8

Force [kN]

Inter-Storey Drift [%]

Force [kN]

Inter-Storey Drift [%]

Figure 18 - Comparison between Force-Inter-storey drift curves Table 1: major experimental data Test d F *M Ntot Name Type of Specimen Method [mm] [kN] [kNm] Specimen with electro-welded lattice girders slab and Nelson connectors +655.09 +1047.2 22 JB-P1 Cyclic 210 around the column, with reinforcement -680.92 -747.75 plates Specimen with electro-welded lattice girders slab and Nelson connectors +666.70 +892.60 20 JB-P2 Cyclic 210 around the column, without -657.60 -760.23 reinforcement plates Specimen with profiled Steel Sheeting +647.12 +760.09 20 slab and Nelson connectors around JB-S1 Cyclic 210 -639.66 -537.59 the column with reinforcement plates Specimen with profiled Steel Sheeting slab and Nelson connectors around +627.19 +887.46 19 JB-S2 Cyclic 175 the column, without reinforcement -634.58 -636.37 plates * + Sagging Moment; - Hogging Moment The temperature distribution into the composite joint subjected to three different timetemperature curves, as depicted in Figure 19, was implemented by the ABAQUS 6.4.1 software (Hibbitt et al., 2000) and a 3D FE model of joint was studied. Figure 20 shows the mean temperature distribution on the slab as a function of the time of fire exposure obtained from the application of the three inputs. It is evident how the application of the natural fire curve produces a lower rise in temperature in the concrete slab than the application both of the ISO 834 curve and parametric curve proposed in EC1 (UNI EN 1991-1-2 2004).
Temperature [°C] 1200 1000 800 600 400 200 0 0 10 20 30 40 50 Time [min] 60 70 80 90 ISO 834 EC1- Annex A OZone V2.2
Temperature [°C] 300 250 200 150 100 50 0 0 5 10 15 20 25 30 35 Time [min] 40 45 50 55 60
Curve ISO 834 EC1-Annex A OZone

Figure 19 - Different time-temperature curves Figure. 20 - Mean temperature distribution on applied as a function of the time of fire the slab as a function of the time of fire exposure exposure As a results, Figure 21 shows that all the steel parts exposed to fire increase their temperature very fast, reaching a very high temperature after only 15 minutes. Conversely, in the concrete and in the steel component with a concrete cover rebars, horizontal plate and vertical plate

passing through the column the temperature does not increase so quickly, remaining near the value of the ambient temperature.

Figure 21 - Temperature distribution in the beam-to-column composite joint with prefabricated slab On this basis, it was possible to incorporate degradation characteristics of the material into the mechanical component model based on the “Strength Reduction Factor-SRF”, which is really a strength retention factor present in the current European Design codes (CEN, Eurocode 3-1-2, 2003). This is basically the residual strength of the materials steel or concrete at a particular temperature relative to its basic yield strength at room temperature. Figure 22 shows the reduction of the design moment capacity of the joint as a function of the time of fire exposure.
Sagging Moment (Ozone)
1400 1200 t=20°C t=15 min t=30 min
-1400 -1200 -1000

Hogging Moment (Ozone)
t=20°C t=15 min t=30 min

M [kNm]

1000 800 600 400 200 0 0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 1

M [kNm]

-800 -600 -400 -200 0 -0.2 -0.6 -1 -1.4 -1.8 -2.2 -2.6 -3

Rotation [rad]

Rotatio [rad]

Figure 22 - Moment-rotation relationship of the joint as a function of the time of fire exposition It is possible to see how the hogging moment capacity reduces to approximately 90-95% of the initial value after 15 minutes of exposure, while it approaches about 20 per cent of the initial value after 30 minutes. The residual resistance quickly decreases owing to the loss of resistance of the component exposed to fire. The degradation in resistance and stiffness is observed both for sagging and hogging bending moment, respectively; nevertheless these numerical results show that the beam-to-column joint is able to carry the internal action that the beams transfer into the joint for a maximum time of 15 minutes without any passive fire protection: this time is enough to evacuate the building after a severe earthquake. CONCLUSIONS The paper presented part of results of a numerical and experimental study aimed at developing performed based design methods, realistically considering the structural performance under seismic-induced fire loading. In particular, the paper discuss the experimental program carried out on steel-concrete composite beam-to-column joints together with numerical simulations regarding the fire behaviour. Experimental and numerical results showed how the joint details influence the beam-column sub-assemblage response. All sub-assemblages exhibited rigid behaviour for the designed composite joints and good performance in terms of resistance, stiffness, energy dissipation and local ductility; as expected plastic hinges developed in beams as a consequence of the capacity design. Moreover, a behaviour factor of about 4 has been observed for the composite frames analysed. As a results, the joint can be used in Ductility Class M structures. Numerical fire simulations have shown that the joint is able to carry the

internal action for a maximum time of 15 minutes without any passive fire protection: this time is enough to quit the building after a severe earthquake. Moreover, joints endowed with prefabricated slab exhibits a better behavior compared to the joint with a composite slab. ACKNOWLEDGEMENTS The experimental programme was made possible by the grant provided under the RELUIS National programme for seismic engineering: Task n.5 - "Development of innovative approaches on the design of steel and steel-concrete composite structures.” The author express their gratitude to Dr: Marco Molinari and the technicians of the Structural Testing Laboratory of Trento University, who skilfully prepared and followed the tests. REFERENCES Bursi O.S., et al. Editors, Final Report PRECIOUS Project Contr. N. RFSCR-03034, 2008. Prefabricated Composite Beam-to-Concrete Filled Tube or Partially Reinforced-ConcreteEncased Column Connections for Severe Seismic and Fire Loadings; ECCS 1986. Recommended Testing Procedure for Assessing the Behaviour of Structural Steel Elements under Cyclic Loads. ECCS Publication n° 45; Ferrario F., Pucinotti R., Bursi O. S., Zandonini R. 2007, Steel-Concrete Composite Full Strength Joints with Concrete Filled Tubes. Part II: Experimental and Numerical Results, Proceeding of 21st Conference of Steel Structures C.T.A., Catania, 1-3 October, pp. 397-404, Dario Flaccovio Editor; Franssen J.-M. 2000. “SAFIR; Non linear software for fire de-sign”. Univ. of Liege; Hibbitt, Karlsson and Sorensen 2000. “ABAQUS User’s manuals”. 1080 Main Street, Pawtucket, R.I. 02860; RELUIS Task n.5: "Development of innovative approaches on the design of steel and steelconcrete composite structures. Unità di Ricerca 8 - Buri O.S., Zandonini R.; SAC 1997, Protocol for Fabrication, Inspection, Testing, and Documentation of Beam-Column Connection Tests and Other Experimental Specimens, report n. SAC /BD-97/02; UNI EN1991-1-2. 2004. “Eurocode 1: Actions on Structures. Part 1-2 : General Actions – Actions on structures exposed to fire”; UNI EN 1992-1-2. 2005. “Eurocode 2: Design of concrete structures. Part 1-2: General rules Structural fire design; UNI EN 1993-1-1. 2005. “Eurocode 3: Design of steel Structures. Part 1: General rules and rules for buildings”; UNI EN 1993-1-8. 2005. “Eurocode 3: Design of steel Structures - Part 1-8: Design of joints”; UNI EN 1993-1-2. 2005. “Eurocode 3: Design of steel Structures. Part 1-2: General rules Structural fire design”; UNI EN 1994-1-1. 2005. “Eurocode 4: Design of composite steel and concrete Structures - Part 1.1: General rules and rules for buildings”; UNI EN 1994-1-2. 2005. “Eurocode 4: Design of composite steel and concrete Structures - Part 1.2: General rules – Structural fire design”; UNI EN 1998-1. 2005. “Eurocode 8: Design of structures for earthquake resistance - Part 1: General rules, seismic actions and rules for buildings”.


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